SEAINT List Archive: December 2010

Monday, December 6, 2010

RE: Eccentric Loads on Welds

Thanks. You confirmed the direction I was headed.

-----Original Message-----
From: Conrad Harrison [mailto:sch.tectonic@bigpond.com]
Sent: Monday, December 06, 2010 3:48 PM
To: seaint@seaint.org
Subject: RE: Eccentric Loads on Welds

Thor,

Treat the welds as line elements and calculate section modulus for. Section
properties of various weld groups can be found in handbooks: circular, two
vertical lines, 2 horizontal lines etc..

Calculate normal and shear forces, then translate into how the code checks
the combination. Here older codes checked transverse and longitudinal
actions on different shear planes in the weld, current codes largely ignore
and check against a single resistance.

Regards
Conrad Harrison
B.Tech (mfg & mech), MIIE, gradTIEAust
mailto:sch.tectonic@bigpond.com
Adelaide
South Australia

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RE: Eccentric Loads on Welds

Thor,

Treat the welds as line elements and calculate section modulus for. Section
properties of various weld groups can be found in handbooks: circular, two
vertical lines, 2 horizontal lines etc..

Regards
Conrad Harrison
B.Tech (mfg & mech), MIIE, gradTIEAust
mailto:sch.tectonic@bigpond.com
Adelaide
South Australia

soil press at square pad diagonal

I have a situation where a square footing can be loaded at 45 degrees to the axis. I was just sitting down to work out the equation to calculate the footing pressure when the resultant is just outside of the kern. At first glance it looked like the algebra was going to take a bit of time. I don’t have a reference that provides this. I certainly can do it, just thought I would ask if there is a reference someone can give or possibly a copy of pages from a reference or if they just have the equations handy. It would save me some time this week. Of course the way work is nowadays, I can’t really say I don’t have the time.

Thanks,

Joe

Joseph R. Grill, PE

Verde Valley Engineering, PLLC

email: VVEng@cableone.net

Register for SEAOSC Wood Design Webinar

SEAOSC Member Doug Thompson, S.E. will present a Webinar on Wood Design: Significant Changes from 2006 to 2009 IBC Chapter 23-Wood on Wednesday, Dec. 8, 2011. Webinar starts at noon PST and ends at 1 p.m. Register at: http://www.seaosc.org/events_detail_announcement.cfm?pk_announcement=27

Eccentric Loads on Welds

I'm looking at the weldment of an aluminum post to a plate. Apart from the
heat affect on the post alloy I am looking at the weld itself. I find that
for the example I am working on , that a simple flange weld is inadequate
and I am considering the calculation for the weld all around the post
section. That requires consideration of eccentric loading but the handbooks
etc only give examples for a plate welded to a plate.

Can I assume that the "plate thickness", 't' can be the dimension of the
post?

Thanks

Thor A. Tandy P.Eng

Victoria, BC
Canada

TIA 222 G

Hello,

My company need to purchase TIA 222 G for a job. Do I need to buy TIA 222G, TIA 222 G1 as well as TIA 222 G2?.

We haven't used TIA 222G so far.

Thanks for the advise.

Regards

Ashraf

TIA TIA-222-G-2
2009-NOV-24 • Active/Current •
Structural Standard for Antenna Supporting Structures and Antennas-Addendum 2

TIA TIA-222-G-1
2007-APR-11 • Active/Current •
Structural Standards for Steel Antenna Towers and Antenna Supporting Structures – Addendum 1

TIA TIA-222-G
2005-AUG-02 • Active/Current •
Structural Standard for Antenna Supporting Structures and Antennas-EFFECTIVE JANUARY 1, 2006; Includes County Listing 07/20/2005 after TIA-222-G standard

Friday, December 3, 2010

RE: PEMB Foundation Uplift

The load combination of 0.6D+W will always give a factor of safety of
1.5 or greater as long as you include the weight of the footing in the D
and thus satisfy a "1.5 factor of safety" though I am not sure the 1.5
factor of safety is codified in the code I design too but maybe it is
and I don't know it.

I don't think it is the intent of the code to exclude the weight of the
footing from the load combination and separating it without factoring it
by 0.6 is just bad algebra. The expression for required weight of
footing should be expressed as '5/3W-D' with D being the dead load less
footing weight and thus the total load would reduce to 5/3W-D+D or 5/3W
which will be greater than 3/2D mentioned.

Using the expression 'W-0.6D' to determine required footing weight or
even 150% of 'W-0.6D' will not satisfy the required load combinations...
at least that is my interpretation.

Donny Harris SE
Los Angeles, CA

Subject: RE: PEMB Foundation Uplift
From: Long Chen <longfchen@gmail.com>
To: seaint@seaint.org

--000e0cd5d020df1427049675ae86
Content-Type: text/plain; charset=ISO-8859-1

Hello, Jeff,

I think that the 1.5 factor is needed. Let's say that the wind load is W
and
the dead load beside the footing is D. To make it simple, let's say that
both are given in positive values. In order to provide an uplifting
safety
factor of 1.5 or greater, we will need to have "D + footing weight" be
greater than 1.5W.

When you design the footing with the net uplift load 0.6D+W, you
actually
get a footing weight of W-0.6D since D and W counteract each other. Then
the
total dead load including the footing is D+(W-0.6D) = W+0.4D, which may
or
may not be larger than 1.5W. Then the safety factor achieved may be
smaller
than 1.5.

But if you design the footing with 1.5 times the net uplift load, you
will
get a footing weight of 1.5 * (W-0.6D) = 1.5W-0.9D. Adding this back to
D
you get a total dead load including footing of D + 1.5W - 0.9D =
1.5W+0.1D.
This value will always be greater than 1.5W since D can only be
positive.
And accordingly you will get a safety factor larger than 1.5 as
required,
regardless how small D is.

Hope that can help.

Regards,
Long

--
Long F. Chen, P.E.
Civil/Structural Engineer
Honolulu, Hawaii

From: Jeff Hedman [mailto:jeff_h@lrpope.com]=20
Sent: Wednesday, December 01, 2010 8:27 AM
To: seaint@seaint.org
Subject: PEMB Foundation Uplift

I designed a foundation for a PEMB. I want to know peoples methods for
designing the footings for uplift. I have typically designed the
footing to weigh as much as the net uplift load (0.6D+(W or 0.7E)). The
calculated net uplift force has a factor of safety of 1/0.6 =3D 1.67 in
=
it
from the reduction of the dead load. I received a plan check comment
that states that with my calculations there is only a factor of safety
of 1. But this is a factor of safety of 1 in regards to the net uplift
force. Should we really be increasing the footing weight to maintain a
factor of safety =3D 1.5 over the net uplift force? To me this is more
like a factor of safety of 1.67*1.5 =3D 2.51. Not to mention we know =
what
the actual weight of the footing will be. This is a commercial building
and the city inspectors are very good in requiring construction matches
the plans, especially foundations. Am I wrong? Do others design PEMB
footings that weigh 1.5 times the net uplift force?

Thanks in advance for your replies

Jeff Hedman

9675ae86--

Re: Beam properties

Looking at a Bethlehem Steel Shapes manual from 1962, looks like the closet match is a 12 I 31.8, standard beam.

On Fri, Dec 3, 2010 at 1:10 PM, Mark@Crownjade.com <mark@crownjade.com> wrote:

I have an existing building with RTU's being added and need to analyze the beams. The simplest of course is a table of properties. The beam dimensions don't fit W, M, S, or HP per AISC 9^th. It is an I-shape with a slightly tapered flange (I didn't get that difference in thickness dimension), d=12-3/16", bf=5", tf=3/8" (by subtraction) and tw=1/4". The beams are fairly light. Does anybody know where to find the properties for these? They are in a 50 to 80 year old Denver, CO building.

Thank You, Mark Benjamin

Crown Jade Design and Engineering, Inc.

(970) 472 2394 www.crownjade.com

Engineered For the Way You Build...

...Designed For the Way You Live

Licensed in AZ, CO, CT, ID, KY, MN, NM, PA, WY... and your state!

P Please do not print this e-mail unless it is completely necessary! Think about the environment.

Confidentiality Statement: This message is solely for the use of the intended recipient. This message is confidential, and may contain legally-privileged information. If you are not the intended recipient, be aware that any review, disclosure, copying, distribution, reliance upon or use of the contents of this message is strictly prohibited. If you have received this message in error, please destroy it and notify me immediately. Thank you.

--
David Topete, SE

RE: Beam properties

Get a copy of the AISC Search Utility for Structural Shapes.

The member may be an S12X25 manufactured by Phoenix Iron Company.

Library: HISTORIC

Type: S

ID = 346

Designation = S 12x25

A = 7.35

D = 12.00

TW = 0.240

BF = 5.000

TF = 0.464

T = 9.882

K = 1.059

W = 25.0

BF_2TF = 5.39

H_TW = 41.2

FY3P = 38

X1 = 2546

X2 = 3769

IX = 182.7

ZX = 34.0

SX = 30.5

RX = 4.99

IY = 7.6

ZY = 5.20

SY = 3.0

RY = 1.02

J = 0.51

CW = 253

a = 35.8

WNO = 14.4

SW = 8.35

QF = 6.36

QW = 17.0

FStr = PH 1915,PH 1923,PH 1929,

From: Mark@Crownjade.com [mailto:mark@crownjade.com]
Sent: Friday, December 03, 2010 12:11 PM
To: seaint@seaint.org
Subject: Beam properties

I have an existing building with RTU’s being added and need to analyze the beams. The simplest of course is a table of properties. The beam dimensions don’t fit W, M, S, or HP per AISC 9^th. It is an I-shape with a slightly tapered flange (I didn’t get that difference in thickness dimension), d=12-3/16”, bf=5”, tf=3/8” (by subtraction) and tw=1/4”. The beams are fairly light. Does anybody know where to find the properties for these? They are in a 50 to 80 year old Denver, CO building.

Thank You, Mark Benjamin

Crown Jade Design and Engineering, Inc.

(970) 472 2394 www.crownjade.com

Engineered For the Way You Build...

...Designed For the Way You Live

Licensed in AZ, CO, CT, ID, KY, MN, NM, PA, WY... and your state!

P Please do not print this e-mail unless it is completely necessary! Think about the environment.

Confidentiality Statement: This message is solely for the use of the intended recipient. This message is confidential, and may contain legally-privileged information. If you are not the intended recipient, be aware that any review, disclosure, copying, distribution, reliance upon or use of the contents of this message is strictly prohibited. If you have received this message in error, please destroy it and notify me immediately. Thank you.

Beam properties

Thank You, Mark Benjamin

Crown Jade Design and Engineering, Inc.

(970) 472 2394 www.crownjade.com

Engineered For the Way You Build...

...Designed For the Way You Live

Licensed in AZ, CO, CT, ID, KY, MN, NM, PA, WY... and your state!

P Please do not print this e-mail unless it is completely necessary! Think about the environment.

Re: PEMB Foundation Uplift

Jeff:

If you have footings and slab that are connected to the pad, you could include some of those in the dead load as well.

Farzin S. Rahbar, SE

Vice President
David C. Weiss Structural Engineer & Associates, Inc.
(818) 227-8040 Ex. 13 Fax: (818) 227-8041

-----Original Message-----
From: Jeff Hedman <jeff_h@lrpope.com>
To: seaint@seaint.org
Sent: Fri, Dec 3, 2010 8:42 am
Subject: RE: PEMB Foundation Uplift

Long,

I agree with you. I was thinking about the same thing last night. If I have 12300 lbs of uplift and 2700 lbs of structure weight, 0.6D+W = 0.6*2700 – 12300 = 10680 lbs of uplift. I would then design the anchor bolts for this force. If I designed the footing to weigh 10680 lbs, my actual factor of safety would be (10680+2700)/12300 = 1.09, probably not necessarily the best idea. So I have decided I will just multiply the footing weight by 0.6 and have even larger footings. Although I am pretty sure I wont see the building rolling down the street in the next wind storm.

Jeff

RE: PEMB Foundation Uplift

I recall a Structural Engineer or Structure magazine article that laid this
out a bit more clearly. I tried to find the article on-line, but have had
limited success. To summarize my memory:

The older codes used a 1.5 overturning / uplift safety factor. Meaning that
your stabilizing dead load had to be 1.5 times the uplift wind load. But,
that was with a 0.9 factor on the dead load. Therefore, the total safety
factor was 1.5 / 0.9 = 1.67.

The new codes use a 0.6 Dead Load factor. Therefore, if you use a 1.0
safety factor with these LC's then you will get a total safety factor of 1.0
/ 0.6 - 1.67.... The same as the older codes.

Here is a simple example using an 11 kip wind uplift and a 2 kip stabilizing
dead load:

Old method:
0.9 (DL + SW) > 1.5 * W
SW reqd > (11*1.5 - 0.9*2) / (0.9) = 16.3 kips

Similarly, using a 1.0 safety factor with the new code LC's would be:

0.6 (DL+SW) > 1.0 *W
SW required > (1.0*11-0.2*0.6) / 0.6 = 16.3 kips

Therefore, you get the same requirement using the old and new codes. No
real changes, just a reconfiguration of where the safety factors are placed
(in the LC rather than in a separate safety factor).

Sincerely,

Josh Plummer

-----Original Message-----

From: Jeff Hedman [mailto:jeff_h@lrpope.com]=20
Sent: Wednesday, December 01, 2010 8:27 AM
To: seaint@seaint.org
Subject: PEMB Foundation Uplift

I designed a foundation for a PEMB. I want to know peoples methods for
designing the footings for uplift. I have typically designed the
footing to weigh as much as the net uplift load (0.6D+(W or 0.7E)). The
calculated net uplift force has a factor of safety of 1/0.6 =3D 1.67 in =
it
from the reduction of the dead load. I received a plan check comment
that states that with my calculations there is only a factor of safety
of 1. But this is a factor of safety of 1 in regards to the net uplift
force. Should we really be increasing the footing weight to maintain a
factor of safety =3D 1.5 over the net uplift force? To me this is more
like a factor of safety of 1.67*1.5 =3D 2.51. Not to mention we know =
what
the actual weight of the footing will be. This is a commercial building
and the city inspectors are very good in requiring construction matches
the plans, especially foundations. Am I wrong? Do others design PEMB
footings that weigh 1.5 times the net uplift force?

Thanks in advance for your replies

Jeff Hedman

RE: PEMB Foundation Uplift

Long,

Jeff

Thursday, December 2, 2010

RE: PEMB Foundation Uplift

Hello, Jeff,

I think that the 1.5 factor is needed. Let's say that the wind load is W and the dead load beside the footing is D. To make it simple, let's say that both are given in positive values. In order to provide an uplifting safety factor of 1.5 or greater, we will need to have "D + footing weight" be greater than 1.5W.

When you design the footing with the net uplift load 0.6D+W, you actually get a footing weight of W-0.6D since D and W counteract each other. Then the total dead load including the footing is D+(W-0.6D) = W+0.4D, which may or may not be larger than 1.5W. Then the safety factor achieved may be smaller than 1.5.

But if you design the footing with 1.5 times the net uplift load, you will get a footing weight of 1.5 * (W-0.6D) = 1.5W-0.9D. Adding this back to D you get a total dead load including footing of D + 1.5W - 0.9D = 1.5W+0.1D. This value will always be greater than 1.5W since D can only be positive. And accordingly you will get a safety factor larger than 1.5 as required, regardless how small D is.

Hope that can help.

Regards,
Long

--
Long F. Chen, P.E.
Civil/Structural Engineer
Honolulu, Hawaii

From: Jeff Hedman [mailto:jeff_h@lrpope.com]=20
Sent: Wednesday, December 01, 2010 8:27 AM
To: seaint@seaint.org
Subject: PEMB Foundation Uplift

I designed a foundation for a PEMB. I want to know peoples methods for
designing the footings for uplift. I have typically designed the
footing to weigh as much as the net uplift load (0.6D+(W or 0.7E)). The
calculated net uplift force has a factor of safety of 1/0.6 =3D 1.67 in =
it
from the reduction of the dead load. I received a plan check comment
that states that with my calculations there is only a factor of safety
of 1. But this is a factor of safety of 1 in regards to the net uplift
force. Should we really be increasing the footing weight to maintain a
factor of safety =3D 1.5 over the net uplift force? To me this is more
like a factor of safety of 1.67*1.5 =3D 2.51. Not to mention we know =
what
the actual weight of the footing will be. This is a commercial building
and the city inspectors are very good in requiring construction matches
the plans, especially foundations. Am I wrong? Do others design PEMB
footings that weigh 1.5 times the net uplift force?

Thanks in advance for your replies

Jeff Hedman

Wednesday, December 1, 2010

RE: IBC 2006/2009 Assumption of flexible diaphragm.

By using the prescriptive criteria of Section 1613.6.1 basically I do not need to run calculations to verify the flexibility of the diaphragm. Per your opinion proprietary prefabricated SWs may not be included under item 3 of this section unless further justifications and calculations are carried out.

David, Thanks very much for your opinion.

Freeman Tang

From: David Topete [mailto:d.topete73@gmail.com]
Sent: Wednesday, December 01, 2010 5:34 PM
To: seaint@seaint.org
Subject: Re: IBC 2006/2009 Assumption of flexible diaphragm.

Again, I would consult the technical documentation of each product considered. For example, the Simpson Steel Strongwall has drift limits. I think it will come down to calculating the deflections of both the horizontal and vertical diaphragms and determining whether it's "flexible' or "rigid."

The reality is that most floors or roofs in residential construction would technically be considered "rigid" because of the diaphragm's aspect ratio, meaning the max deflection at midspan usually will not be greater than 2.5 times the deflection of the shearwalls or moment frames supporting them.

HTH.

On Wed, Dec 1, 2010 at 4:19 PM, Freeman Tang <ftang@gouvisgroup.com> wrote:

David,

Thanks for the quick reply. The R factor for Hardy Panels is 6.5. The R factor takes care of the ductility issue, but not the SW's stiffness relative to the diaphragm's stiffness. If a horizontal wood diaphragm is supported by moment frames, it will behave more rigid than if it is supported by long concrete SWs. So the flexibility of a diaphragm is relative to the stiffness of the vertical resisting elements supporting it. If the diaphragm is supported by slender prefabricated SWs, it will behave more rigid than if its is supported by long wood SWs.

In the commentary of IBC 2006 item 3 limits this allowance of flexible diaphragm assumption to light-framed wood structures to those seismic-force-resisting system that traditionally have been designed assuming flexible diaphragm. Does the "traditionally" include prefabricated SWs? This is my doubt.

Freeman

From: David Topete [mailto:d.topete73@gmail.com]
Sent: Wednesday, December 01, 2010 2:50 PM
To: seaint@seaint.org
Subject: Re: IBC 2006/2009 Assumption of flexible diaphragm.

I would say they would be considered the same, unless their respective documentation or testing report or ICC ESR indicates the use of an alternate response modification factor, R. I recall Hardy Frames were designed as steel braced frames and indicated an R factor of 4 (or 4.5). Off hand, I don't remember what their Hardy Panels were to be designed for...

On Wed, Dec 1, 2010 at 2:09 PM, Freeman Tang <ftang@gouvisgroup.com> wrote:

Hi list,

Per IBC 2006/2009 of section 1613.6.1 Assumption of flexible diaphragm, item 3 stipulates that "Vertical elements of the ………..with wood structural panels for shear resistance or steel sheets". My question is that will proprietary prefabricated wood panels and steel panels be considered as the wood structural panels and steel sheets under item 3 of this section?

I believe that the wood structural panels and steel sheets under this section refer to the conventional wood shear walls and sheet steel shear panels, but not including the proprietary SWs. Do you have different opinions on this? Any comments and suggestions will be appreciated.

Freeman Tang

--
David Topete, SE

Re: IBC 2006/2009 Assumption of flexible diaphragm.

HTH.

On Wed, Dec 1, 2010 at 4:19 PM, Freeman Tang <ftang@gouvisgroup.com> wrote:

David,

Thanks for the quick reply. The R factor for Hardy Panels is 6.5. The R factor takes care of the ductility issue, but not the SW's stiffness relative to the diaphragm's stiffness. If a horizontal wood diaphragm is supported by moment frames, it will behave more rigid than if it is supported by long concrete SWs. So the flexibility of a diaphragm is relative to the stiffness of the vertical resisting elements supporting it. If the diaphragm is supported by slender prefabricated SWs, it will behave more rigid than if its is supported by long wood SWs.

In the commentary of IBC 2006 item 3 limits this allowance of flexible diaphragm assumption to light-framed wood structures to those seismic-force-resisting system that traditionally have been designed assuming flexible diaphragm. Does the "traditionally" include prefabricated SWs? This is my doubt.

Freeman

From: David Topete [mailto:d.topete73@gmail.com]
Sent: Wednesday, December 01, 2010 2:50 PM
To: seaint@seaint.org
Subject: Re: IBC 2006/2009 Assumption of flexible diaphragm.

I would say they would be considered the same, unless their respective documentation or testing report or ICC ESR indicates the use of an alternate response modification factor, R. I recall Hardy Frames were designed as steel braced frames and indicated an R factor of 4 (or 4.5). Off hand, I don't remember what their Hardy Panels were to be designed for...

On Wed, Dec 1, 2010 at 2:09 PM, Freeman Tang <ftang@gouvisgroup.com> wrote:

Hi list,

Per IBC 2006/2009 of section 1613.6.1 Assumption of flexible diaphragm, item 3 stipulates that "Vertical elements of the ………..with wood structural panels for shear resistance or steel sheets". My question is that will proprietary prefabricated wood panels and steel panels be considered as the wood structural panels and steel sheets under item 3 of this section?

I believe that the wood structural panels and steel sheets under this section refer to the conventional wood shear walls and sheet steel shear panels, but not including the proprietary SWs. Do you have different opinions on this? Any comments and suggestions will be appreciated.

Freeman Tang

--
David Topete, SE

--
David Topete, SE

RE: PEMB Foundation Uplift

The footing weight is D, not 0.6D + W

Jim Lutz, PE, SE
Senior Structural Engineer
BHC Consultants, LLC

1601 5^th Avenue, Suite 500
Seattle, WA 98101
Ph: 206-505-3400
Fax: 206-505-3406
www.bhcconsultants.com

This email and all attachments are confidential. For further information about emails sent to or from BHC Consultants or if you have received this email in error, please refer to http://www.bhcconsultants.com/Email+Disclaimer+/default.aspx

From: Jeff Hedman [mailto:jeff_h@lrpope.com]
Sent: Wednesday, December 01, 2010 9:37 AM
To: seaint@seaint.org
Subject: RE: PEMB Foundation Uplift

Jim,

To clarify, you are saying that you agree with me that the footing weight only needs to be equal to net uplift calculated from 0.6D+W? And that increasing the footing weight 50% to obtain a 1.5 factor of safety over 0.6D+W would be excessive and not required?

Jeff Hedman , S.E.

email logo

L.R. Pope Engineering, Inc.

1240 East 100 South Suite # 15B

St. George, Utah 84790

Office: 435-628-1676

Fax: 435-628-1788

email: jeff_h@lrpope.com

This email message, including any attachments, is for the sole use of the intended recipient(s) and may contain confidential and privileged information. Any unauthorized review, use, disclosure or distribution of this information is prohibited, and may be punishable by law. If you have received this electronic mail transmission in error, please notify the sender by reply e-mail or by telephone and destroy all contents of the original message on all computers or hard copies.

RE: IBC 2006/2009 Assumption of flexible diaphragm.

David,

Freeman

From: David Topete [mailto:d.topete73@gmail.com]
Sent: Wednesday, December 01, 2010 2:50 PM
To: seaint@seaint.org
Subject: Re: IBC 2006/2009 Assumption of flexible diaphragm.

On Wed, Dec 1, 2010 at 2:09 PM, Freeman Tang <ftang@gouvisgroup.com> wrote:

Hi list,

Freeman Tang

--
David Topete, SE

Re: Beam clamps for piping braces

Drew,

You might look at the Power-Strut catalog (see web address below).

http://www.atcelectrical.com/literature/pdf/PSN_SEISMIC.pdf

Thomas Hunt, S.E.
Fluor

From: Drew Morris <dmorris@bbfm.com>
To: SEAINT <seaint@seaint.org>
Date: 11/30/2010 04:13 PM
Subject: Beam clamps for piping braces

Does anyone know if B-Line or Unistrut or others makes a clamp that attaches to the bottom flange of a W beam and is tested for lateral loads and can be used to brace piping / ducts / etc?
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Re: IBC 2006/2009 Assumption of flexible diaphragm.

On Wed, Dec 1, 2010 at 2:09 PM, Freeman Tang <ftang@gouvisgroup.com> wrote:

Hi list,

Per IBC 2006/2009 of section 1613.6.1 Assumption of flexible diaphragm, item 3 stipulates that "Vertical elements of the ………..with wood structural panels for shear resistance or steel sheets". My question is that will proprietary prefabricated wood panels and steel panels be considered as the wood structural panels and steel sheets under item 3 of this section?

I believe that the wood structural panels and steel sheets under this section refer to the conventional wood shear walls and sheet steel shear panels, but not including the proprietary SWs. Do you have different opinions on this? Any comments and suggestions will be appreciated.

Freeman Tang

--
David Topete, SE

IBC 2006/2009 Assumption of flexible diaphragm.

Hi list,

Freeman Tang

RE: PEMB Foundation Uplift

Jim,

Jeff Hedman , S.E.

email logo

L.R. Pope Engineering, Inc.

1240 East 100 South Suite # 15B

St. George, Utah 84790

Office: 435-628-1676

Fax: 435-628-1788

email: jeff_h@lrpope.com

RE: PEMB Foundation Uplift

I think the reason the net uplift equation factor on the dead load was changed to .60 instead from .90 is precisely to guarantee a 1.5 uplift safety factor. Adding another 1.5 safety factor on top of this should not be required.

Jim Lutz, PE, SE
Senior Structural Engineer
BHC Consultants, LLC

From: Jeff Hedman [mailto:jeff_h@lrpope.com]
Sent: Wednesday, December 01, 2010 8:27 AM
To: seaint@seaint.org
Subject: PEMB Foundation Uplift

I designed a foundation for a PEMB. I want to know peoples methods for designing the footings for uplift. I have typically designed the footing to weigh as much as the net uplift load (0.6D+(W or 0.7E)). The calculated net uplift force has a factor of safety of 1/0.6 = 1.67 in it from the reduction of the dead load. I received a plan check comment that states that with my calculations there is only a factor of safety of 1. But this is a factor of safety of 1 in regards to the net uplift force. Should we really be increasing the footing weight to maintain a factor of safety = 1.5 over the net uplift force? To me this is more like a factor of safety of 1.67*1.5 = 2.51. Not to mention we know what the actual weight of the footing will be. This is a commercial building and the city inspectors are very good in requiring construction matches the plans, especially foundations. Am I wrong? Do others design PEMB footings that weigh 1.5 times the net uplift force?

Thanks in advance for your replies

Jeff Hedman

PEMB Foundation Uplift

I designed a foundation for a PEMB. I want to know peoples methods for designing the footings for uplift. I have typically designed the footing to weigh as much as the net uplift load (0.6D+(W or 0.7E)). The calculated net uplift force has a factor of safety of 1/0.6 = 1.67 in it from the reduction of the dead load. I received a plan check comment that states that with my calculations there is only a factor of safety of 1. But this is a factor of safety of 1 in regards to the net uplift force. Should we really be increasing the footing weight to maintain a factor of safety = 1.5 over the net uplift force? To me this is more like a factor of safety of 1.67*1.5 = 2.51. Not to mention we know what the actual weight of the footing will be. This is a commercial building and the city inspectors are very good in requiring construction matches the plans, especially foundations. Am I wrong? Do others design PEMB footings that weigh 1.5 times the net uplift force?

Thanks in advance for your replies

Jeff Hedman